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Volume 29 No.1 April 2016
SESOC Journal
73
DISPLACEMENT-BASED RC COLUMN
ASSESSMENT FOR A CASE STUDY
INTERWAR BUILDING
Kevin Q. Walsh 1,2, Dmytro Y. Dizhur 1,3, Peter Liu 3, Mostafa Masoudi 3, Jason M. Ingham 1
ABSTRACT:
Following the devastating 1931 Hawke’s Bay earthquake, commercial buildings in the Hawke’s Bay, New
Zealand region were rebuilt in mostly homogenous structural and architectural styles. Most were constructed
of reinforced concrete (RC) two-way space frames in the Art Deco aesthetic popularised during the interwar
time period. Although most Art Deco RC columns in Hawke’s Bay have generally ductile detailing for their time
period of construction, they are nonetheless often expected to be brittle, earthquake-prone components based
on strength-based seismic assessments. The reported case study was intended to provide a displacement-
based example for undertaking a seismic assessment of Art Deco RC columns while appropriately accounting
for regional seismicity, material properties, building component interaction, column geometry, and reinforcement
detailing, as a resource for professional structural engineers tasked with seismic assessments and retrofit
designs for similar buildings.
PAPER CLASS & TYPE: GENERAL REFEREED
1 Department of Civil and Environmental Engineering,
University of Auckland, New Zealand
2 Frost Engineering and Consulting, Mishawaka, Indiana, United States
3 EQ STRUC Limited, Auckland, New Zealand
INTRODUCTION
Art Deco buildings are of immense value to the cultural
and civic heritage of the Hawke’s Bay community.
However, a lack of understanding of the expected
performance of these buildings in an earthquake
threatens their continued utility. Past engineering
assessments of these ostensibly brittle reinforced
concrete (RC) low-rise buildings have resulted in
predictions of generally poor seismic performance
(e.g., van de Vorstenbosch et al. 2002), contrary to
the empirical evidence from the 1931 Hawke’s Bay
earthquake (Mitchell 1931; Brodie and Harris 1933) and
empirical evidence from other historical earthquakes
in New Zealand (Dowrick and Rhoades 2000). These
buildings are now threatened with forced vacancy or
demolition by legislation (New Zealand Parliament 2004,
2005) dependent on their estimated seismic capacities.
As a result, other researchers have called for more
sophisticated studies into the seismic capacities of Art
Deco buildings in Hawke’s Bay (Dowrick 2006).
Walsh et al. (2014) carried out a geometric study of
Hawke’s Bay’s Art Deco buildings and performed a
pushover capacity assessment of several RC ground
storey columns within a representative sample of Art
Deco buildings. Compared to the buildings considered by
Walsh et al. (2014), the Information Management Services
(IMS) building in Hastings has a relatively low ratio of the
sum of RC column and wall cross-sectional areas on
the ground floor to the total building footprint area, or
“structural footprint” ratio, of only 0.6%. Furthermore and
as will be demonstrated herein, the typical ground storey
column of the IMS Hastings building was estimated to
have a relatively high ratio of shear demand at plastic
hinging to nominal shear capacity, Vp / Vn, and would be
deemed in accordance with the criteria of the American
Society of Civil Engineers (ASCE 2014) to be likely to
experience an undesirable shear-controlled failure under
lateral loading. However, despite these disadvantages,
this particular building survived the 1931 Hawke’s
Bay earthquake with only superficial damage, making
it an appropriate case study building in the following
displacement-based assessment.
THE CASE STUDY BUILDING
The IMS Hastings building was constructed by the
Hawke’s Bay Farmers Co-operative Association (HBFCA)
in 1929 to replace the building on the same site that had
been destroyed by a fire. It was designed by Wellington
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architect Edmund Anscombe & Associates and survived
the 3 February 1931 Hawke’s Bay earthquake with little
observable structural damage (NZHPT 2013) despite
the high local intensities (>MM9) and in contrast to
the extensive damage to many neighbouring buildings
caused by the earthquake (Dowrick 1998). The IMS
Hastings building is three storeys in height above grade,
and it has a basement that comprises approximately half
of the building footprint which is approximately
30 m x 36 m. The building’s support structure is
comprised of multiple bays of RC columns typically
spaced at 6 m x 6 m. The columns are octagonal in
cross-section with an effectively equivalent circular
cross-sectional diameter of 447 mm at the ground storey.
The tops of the columns taper outward into a thickened
slab haunch (i.e., “drop slab” construction) supporting
two-way spanning RC floor slabs. RC spandrel beams
extend around the exterior of the building at each level.
In addition to unreinforced clay brick masonry (URM)
infill walls on the exterior of the building, interior URM
walls exist in multiple locations throughout the building in
addition to two RC lift shafts. An “uppercroft” annex was
added to the roof with access from the second storey
over a portion of the building footprint. Photographs
of the typical exterior and interior arrangement of
the building are included in Figure 1, and additional
illustrations and details are included in Appendix A.
(a) Southeast elevation of the building
(b) Northwest elevation of the building showing URM infill walls
(c) First storey internal column arrangement
(d) Ground storey internal column arrangment
Figure 1. Photographs of the IMS Hastings building
REGIONAL SEISMICITY AND PERFORMANCE
CRITERIA
The aggregated hazard factor, Z, in Hastings of 0.39
is three times the aggregated seismic hazard factor for
Auckland (New Zealand’s largest city) and nearly equal to
the aggregated seismic hazard factor in Wellington (New
Zealand’s capital city with the highest seismic hazard in
the country among major urban centres) (NZS 2004).
Short-period (SDS at a period of 0.2 sec) and long-period
(SD1 at a period of 1.0 sec) spectral accelerations for
the design basis earthquake (DBE) and relative levels of
seismicity for the Hastings building site assuming shallow
subsoils (site subsoil class C) are as follows:
• SDS = 1.14 g; and
• SD1 = 0.46 g.
These design spectral accelerations result in Hastings
being considered a region of high seismicity relative to
an international scale (ASCE 2014). Design loadings
standards in New Zealand (NZS 2002) prescribe that
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buildings subjected to DBE actions be designed for
“avoidance of collapse of the structural system… or
parts of the structure… representing a hazard to human
life inside and outside the structure… [and] avoidance
of damage to non-structural systems necessary for…
evacuation.” In accordance with the seismic assessment
guidelines published by the New Zealand Society for
Earthquake Engineering (NZSEE 2006), the emphasised
performance level considered in the case study
assessment discussed herein is the ultimate limit state
(ULS), which is theoretically equivalent to the life safety
(LS) performance level considered by ASCE (2014).
The GNS geological maps (Lee et al. 2011) classify the
IMS Hastings building site as resting on beach deposits
which consist of sand, gravel, silt and mud on modern
coastal plains and lake margins. However, shallow core
samples (< 700 mm) taken from below the basement
slab were determined to indicate better soil conditions
than the classification prescribed in the geological maps.
Site subsoil class C (shallow subsoils) was assumed
in the seismic assessment based on these qualitative
observations, but the analysis was composed to
accommodate site subsoil class D (deep subsoils) as well
should future geotechnical investigations determine that
such subsoil conditions exist.
ASSESSMENT ASSUMPTIONS AND GENERAL
OBSERVATIONS
The DBE assumed for the case study assessment of the
IMS Hastings building has an average return period of
500 years, which is appropriate for a normal building of
importance level 2 (NZS 2002). The building was assessed
for a future 50-year working life. The summary of design
loads and general seismic factors are presented in Table 1.
Table 1. Design (assessment) loads for the IMS Hastings building
Type of load Load magnitude
Corrugated metal roofing
& purlins
0.15 kPa
RC self-weight 24 kN/m3
URM self-weight 17 kN/m3
Suspended ceiling 0.15 kPa
Services and lighting 0.2 kPa
Floor loading (office &
retail)
3.0 kPa, 2.7 kN
Corridors and stairway 4.0 kPa, 4.5 kN
Roof (accessible) 1.5 KPa, 1.8 kN
Hastings District Council holds drawings for the building
that are mostly architectural and consist of plans of
interior fit-out and veranda extensions constructed in
the 1970s. Structural drawings of the original building
construction were not available from the council nor from
the building owner. Instead, invasive investigations were
coupled with non-invasive ferro scanning to determine
the original construction details of the support columns
and floor slabs (see Appendix A). It was observed that
the building was in excellent condition for its age. At the
time of inspection, there were no visible signs to suggest
deterioration of the concrete or the reinforcing steel.
The frame-wall connection mechanism was identified
by invasive investigation at one of the URM infill walls in
the first storey where twisted wire loops were observed
between the concrete members and the mortar joints of
the URM [see Figure 2(a)]. The URM infill walls along the
building perimeter in the first and second storeys were
measured to be at least three wythes thick. The URM
infill walls along the west perimeter of the building were
found to feature cavity construction (i.e., URM wythes
separated by air gaps) bounded by RC spandrels on the
top and bottom. At the first storey column where invasive
investigation was carried out to expose the reinforcing
steel, it was found that the anchored slab haunch area
was detailed with multiple layers of steel reinforcement
and diagonal ties anchored into the octagonal column.
Additional steel reinforcement detailing was identified
without the aid of original structural construction plans
through the removal of concrete cover as exhibited by the
photographs in Figure 2.
The foundation of the IMS Hastings building was
understood to consist of RC grade beams extending
between spread footings beneath the RC columns and
topped with a 200+ mm thick RC slab. The foundation
arrangement was assumed based on a review of other
buildings designed by the same architect and 700 mm
deep concrete cores that were drilled into the basement
slab. These traits are consistent with the observations of
Brodie and Harris (1933) for contemporary RC building
foundations in Hawke’s Bay. The RC basement perimeter
walls of the IMS Hastings building are fully buried and
were estimated to be at least 280 mm thick.
The octagonal columns gradually reduce in cross-
sectional area from the basement to the second storey.
From invasive investigation, it was determined that the
columns were reinforced with eight main longitudinal bars
and spiral reinforcing steel. The columns taper outward
within each storey at approximately 80% of the column
height above the floor slab. From invasive investigation,
it was found that the reinforcement generally follows the
shape of the taper into the floor slab haunch area.
Permanent dead loads
Superimposed dead loads
Live loads
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The depth of the slab haunch at each column is
approximately 275 mm and the slab haunch is reinforced
with at least two layers of reinforcing steel and a number
of diagonal reinforcing bars from the column. The floor
slab was investigated on-site using a rebar scope and
was estimated to be reinforced with two layers of two-
way spanning reinforcing steel. The thickness of the slab
was measured as 175 mm at one exposed location in
the first storey, and this dimension was confirmed by core
sampling in other locations.
The spandrel beams were observed to be reinforced
with larger diameter horizontal reinforcing steel bars
at the top and bottom and smaller diameter vertical
reinforcing bars. RC core walls were determined to be
detailed with a single layer of steel reinforcement. Several
other internal walls in the building were found to have
been constructed of URM with reddish clay bricks and
strong cement-based mortar. Generally, both internal and
external infill URM was found to be in excellent condition
and protected from weathering with a layer of rendering
plaster (in the case of the external surfaces). Additional
findings and specific reinforcement details from the
forensic investigation of the IMS Hastings building are
illustrated in Appendix A.
(a) Twisted metal ties (b) Multilayered reinforcement
embedded between the exposed an area of broken slab
concrete and the URM haunch (ground storey)
infill
(c) Multilayered (d) Parapet profile along the building
reinforcement in the first perimeter with parapet partially
storey slab haunch and in deconstructed for inspection
the base of the column
Figure 2. General observations made
during the invasive investigation
MEASURED MATERIAL PROPERTIES
Concrete core and steel reinforcement samples were
extracted from multiple locations and tested in a
laboratory. The results of the steel reinforcement testing
are summarised in Table 2 and the results of compression
tests on 95 mm diameter concrete core samples are
summarised in Table 3.
Table 2. Summary of test results of the steel reinforcement
samples
Characteristic # tests Mean Stan. dev.
Yield stress, fy (MPa) 11 270 41
Ultimate stress (MPa) 13 399 50
Strain at ultimate stress
(mm/mm)
11 0.1309 0.0677
Table 3. Summary of test results of the concrete core samples
Characteristic # tests Mean Stan. dev.
Max compressive
stress, f'co (MPa)
5 30.28 11.19
Elastic compression
modulus, Ec (GPa)
5 26.53 5.36
STRUCTURAL FOOTPRINT RATIO
Information pertaining to the ratio of the sum of RC
column and wall cross-sectional areas on the ground
floor to the total building footprint area, or structural
footprint ratio, was determined by Walsh et al. (2014) for
a sample of Hawke’s Bay Art Deco buildings. Of the 1083
m2 footprint of the IMS Hastings building, approximately
0.6% of the footprint was comprised of RC columns
(excluding the infill masonry area), which was well below
the average structural footprint ratio of 1.5% for the Walsh
et al. (2014) sample group. Structural footprint ratios
of 0.6 and 0.9% were recommended as minimums by
Glogau (1980) for two-storey and three-storey buildings,
respectively, based on a study of the seismic performance
of low-rise RC buildings of limited ductility in Japan.
However, in a study specific to low-rise RC building
performance in the 1931 Hawke’s Bay earthquake, van
de Vorstenbosch et al. (2002) determined that open
moment-resisting frame systems with structural footprint
ratios of approximately 0.4% or greater performed well,
and infilled moment-resisting frames performed well with
structural footprint ratios as low as 0.3%, excluding the
infill masonry area. Hence, the practicing engineer could
anticipate that a relatively high structural footprint ratio
would limit ULS drift demands imposed on the columns
by the DBE.
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ESTIMATING COLUMN DISPLACEMENT
DEMANDS BY NONLINEAR TIME-HISTORY
ANALYSIS
The nonlinear time-history analysis (NLTHA) used to
estimate the displacement demands on the columns
of the case study building was comprised of a three-
dimensional nonlinear computer model of the building
created in a finite/frame element software program
(SAP2000TM in this case, although other software
programs are also suitable) subjected to a suite of
earthquake actions prescribed for the North Island of New
Zealand (Oyarzo-Vera et al. 2012). Simulated behaviour
amongst all of the structural and pseudo-structural
elements of the building was monitored. The spandrel
was modelled as a nonlinear shell element with different
reinforcement detailing in the longitudinal and vertical
directions. The slab haunches and floor slabs were
represented in the model as nonlinear shell elements with
two layers of two-way spanning reinforcing mesh. It is
generally recommended that infill walls be modelled as
equivalent struts in accordance with NZSEE (2006) and
ASCE (2014) criteria.
For performing a NLTHA, NZS (2004) prescribes that a
minimum of three relevant earthquake records be applied
to the building model and the worst-case scenario be
assumed to govern the demands. Three ground-motion
records were chosen from the list of historical records
deemed appropriate for Hawke’s Bay by Oyarzo-Vera et
al. (2012) whose suggested records were differentiated
for different soil types. In order to ensure that the analysis
could be easily modified if a lower soil quality were
determined based on future soil borings (i.e., change from
site subclass C to D), records were chosen for this model
that were common to both site subsoil class records
suites: El Centro 1940, Tabas 1978, and Hokkaido 2003.
As per NZS (2004), scaling of the ground acceleration
records was required such that the spectral accelerations
of the family of records closely matched the target
demand spectrum within the range of interest of 0.4T1 to
1.3T1 , where T1 represents the first mode (fundamental)
period for the IMS Hastings building of 0.37 sec as
estimated from an elastic spectral response analysis
performed previous to the NLTHA. The effect of scaling
the spectral responses for the three records chosen is
represented graphically in Figure 3.
Figure 3. Earthquake records scaled to the ULS DBE target
spectrum for Hastings site subsoil class C
The time-history scaling factors (k1 x k2) associated
with the El Centro, Tabas, and Hokkaido records,
respectively, were determined to be 1.50, 0.49, and 1.60
for the ULS DBE (i.e., 100%NBS) demands assuming
a structural displacement ductility factor, μ, of 1.25,
which is conservatively low for most Hawke’s Bay Art
Deco buildings (Walsh et al. 2014). The raw horizontal
accelerations were scaled by these factors and input
into the building model as time-history functions. Each
earthquake record was considered in two separate load
cases so that the H1 and H2 directional accelerations
were applied in both orthogonal directions of the building.
The selected pairs of ground motion acceleration records
were applied as the input for nonlinear direct integration
time-history analyses (per NZS 2004), as opposed to
the accelerations being applied to a modal time-history
analysis, although the latter is less computationally
intensive. Geometric nonlinearity parameters were set
to P-Delta. Cracked stiffness of concrete cross-sections
was assumed as 0.4EcAg for the perimeter walls, and
measured material properties obtained from testing were
utilised in the NLTHA models whenever possible with
the intention of estimating displacement demands as
accurately as possible.
When subjected to lateral loads, the displaced profile of
the building model indicated a torsional response with
the peak displacement occurring at opposing diagonal
corners of the building. An exaggerated image of the
building’s displacement at the peak acceleration when
subjected to the scaled Tabas earthquake at 100%NBS
ground accelerations is shown in Figure 4. The torsional
response of the building was mostly attributed to the
continuous shop openings along the street frontage.
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a) 3D model with frame and shell elements
b) Exaggerated displaced shape with Tabas
record peak ground acceleration applied
Figure 4. Modelled response of the IMS Hastings building
Based explicitly on strength and lateral displacement
capacity, the critical columns of the building were
determined to be the ground storey columns in the main
retail area at the street corner where the frame lacks
additional support from the URM infill at the perimeter.
As expected based on the building’s configuration, the
results of the computer-aided analysis indicated that the
ground storey columns in the northwest portion of the
building (see Figure 5) were less critical in comparison to
other ground storey columns due to the additional lateral
capacity from the perimeter walls and lift cores. The in-
plane stresses in the floor slab and slab haunch areas also
appeared to be within the capacity of these RC elements
as determined from simple hand calculations (NZS 2006).
The results of the NLTHA strength assessment indicated
that some columns of the IMS Hastings building were
likely to be over-stressed by the DBE. However such
strength exceedances do not necessarily indicate
occupant life safety-threatening failures at the ULS. If the
columns do not drift significantly, then axial failure of the
columns and partial or total collapse of RC frames within
the building may still be averted, especially considering
how redundant the frames of most Hawke’s Bay Art Deco
buildings, including the IMS Hastings building, appear to
be. Assuming that the slab and slab haunches remain
elastic based on the results of the strength-to-demand
ratios computed from the model, hinges were predicted to
occur at the top and bottom of the columns in accordance
with the hinge properties recommended by FEMA (2000)
and ASCE (2014). Figure 5 includes a floor plan of the
building at which inter-storey drifts were tracked in each of
the time-history scenarios for the column locations labelled
A–E. A summary of the maximum inter-storey drifts from
the three time-history cases considered is included in
Table 4. The torsional response of the building is clearly
identifiable when comparing the storey drift values at
column location A relative to the other tracked column
locations within the building.
Figure 5. Ground storey plan showing column positions
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Table 4. Summary of inter-storey drifts by earthquake ground motion record and column location
Column
location
El Centro Hokkaido Tabas
Ground
storey
1st storey 2nd storey Ground
storey
1st storey 2nd storey Ground
storey
1st storey 2nd storey
A 1.00% 1.31% 1.24% 0.99% 1.33% 1.30% 0.94% 1.22% 1.16%
B 0.74% 1.08% 1.16% 0.69% 1.04% 1.14% 0.61% 0.93% 1.00%
C 0.18% 0.40% 0.53% 0.21% 0.47% 0.59% 0.29% 0.63% 0.78%
D 0.27% 0.59% 0.78% 0.27% 0.60% 0.78% 0.26% 0.61% 0.81%
E 0.51% 0.84% 1.04% 0.52% 0.87% 1.07% 0.49% 0.83% 0.99%
ESTIMATING COLUMN DISPLACEMENT
CAPACITIES BY NONLINEAR PUSHOVER
ANALYSIS
The load and displacement capacity provided by the
geometry and the steel reinforcement detailing in the
ground floor columns of interwar RC buildings is an
especially critical consideration for the assessment of
collapse prevention and expected damage concentration
(Brodie and Harris 1933; Dowrick 1998). For purposes
of developing the worked example, the assessment of
column E on the ground storey (see Figure 5, Table 4, and
Appendix A) to determine its displacement capacity was
recorded in detail as reported herein.
Assumed material properties
In lieu of material test results, NZSEE (2006, sect. 7.1.1)
and TNZ (2004, chap. 6) can be consulted for assumed
material properties for analysis. For the column in this
worked example, however, several material test results
were measured directly (see the mean values listed in
Table 2 and Table 3). Longitudinal (notated as “long”
herein) and transverse (notated as “trans” herein) steel
reinforcement material properties were assumed to be
equivalent. The elastic modulus of the steel reinforcement,
Es , was assumed to be equal to 200,000 MPa, with the
yield strain, İy , being determined as the ratio of the mean
yield stress, fy , listed in Table 2, and the elastic modulus,
Es , resulting in İy = 0.00135.
The concrete unconfined ultimate compressive strength,
f'co , was assumed to be equal to the measured mean
value of 30.3 MPa as listed in Table 3. Concrete elastic
compression modulus, Ec , was assumed to equal the
measured mean of 26.5 GPa as listed in Table 3. Note
that this value is nearly equivalent to the value estimated
by following the recommendations of NZS (2006, sect.
5.2.3) that “for normal density concrete, Ec may be
considered as (3320¥f'co + 6900) MPa.” In accordance
with the recommendation of NZS (2006, sect. C6.9.1),
the material shear modulus for concrete was estimated as
G = 0.4Ec = 10.6 GPa.
Column geometry, reinforcement detailing, and axial
loads
Invasive investigations were coupled with non-invasive
ferro scanning in order to determine the column
reinforcement details (see Appendix A). The geometry and
reinforcement detailing of the ground floor columns of the
IMS Hastings building were determined to be as follows:
• Octagonal column geometry with each of the eight
sides being 185 mm long, effectively equivalent to a
column with circular cross-sectional diameter D = 447
mm and gross cross-sectional area of Ag = 0.157 m2;
• Spiral transverse reinforcement bar of diameter dtrans
= 6.4 mm vertically spaced centre-to-centre at s = 90
mm pitch;
• Eight longitudinal reinforcement bars of diameter
dlong = 28.6 mm with 1070 mm lap splice spaced
approximately equivalently around the inside of the
spiral transverse reinforcement (i.e., one longitudinal
bar located every 45 degrees around the circular
cross-section); and
• 75 mm average concrete clear cover from the outside
edge of the column to the outside edge of the spiral
transverse reinforcement bar.
The ground storey column clear-height to be deformed
in double curvature with assumed flexurally rigid end
restraints, L , was measured as 3930 mm (see Appendix
A). If the direction of lateral loading being considered on
a given column is in-plane with partial-height masonry
infill (as is common on the exterior of Hawke’s Bay Art
Deco buildings per Walsh et al. 2014), then the column
height in this direction of loading should be assumed to
be the clear-storey height minus the infill wall height. Such
a scenario may represent a “short column” vulnerability
(NZSEE 2014), and the column failure mechanism is more
likely to be shear-controlled in this direction of loading.
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In the case of the considered column E (see Figure 5),
however, no mid-height restraint such as masonry infill
was present.
For the case study ground floor column E (see Figure 5),
the concentrated axial load, N , was determined from the
analytical results of the computer-aided model. All self-
weight and imposed gravity loads (see Table 1) above the
column resulted in a total axial compression load of 990
kN. The minimum and maximum total axial compression
load conditions from any of the earthquake loading
scenarios were 945 kN and 1053 kN, respectively. Note
that the bandwidth of axial load extremes was determined
to be relatively small for the considered interior column
(only approximately +/- 5% from the gravity axial load
of 990 kN). Exterior columns are likely to experience
relatively larger bandwidths of axial loads effectuated by
frame action during earthquake loading than are interior
columns, and these axial loads can be estimated from
the results of the computer-aided NLTHA. In this case,
the engineer should consider that the minimum shear
strength of the column will occur when the column is
under the minimum axial compression load, and the
drift ratio at axial load failure may be governed by the
maximum axial compression load.
Idealised backbone pushover model
An idealised backbone model defining the key damage
states of an existing reinforced concrete column under
lateral loading (i.e., flexural yielding, shear failure, and
axial load failure) was developed by Elwood and Moehle
(2006). The nominal moment capacity of the column
cross-section, Mn , was determined to be equal to 277
kN-m for the case study column assuming that the entire
cross-section of the column was intact (i.e., no spalling
of the cover concrete), the minimum axial load of 945 kN
was applied, and the strength of the steel reinforcement
was equal to the measured mean yield stress, fy , of 270
MPa as listed in Table 2 . In this instance, the plastic
moment capacity was determined by transforming the
column into an equivalent rectangular column using the
empirical method proposed by Whitney (1942), with
the result confirmed by using the sectional analysis
program Response-2000TM (Bentz 2000). Accounting
for overstrength caused by strain hardening in the steel
reinforcement by assuming that the strength of the steel
reinforcement was equal to the measured mean ultimate
stress of 399 MPa as listed in Table 2, the probable
moment capacity of the column-cross section, Mp ,
was determined to be equal to 342 kN-m for the case
study column, representing an overstrength ratio of
approximately 1.23. For validation, NZSEE (2006, sect.
7.1.1) notes that “the ratio of overstrength in flexure to
nominal flexural strength… can be taken as 1.25.”
According to Elwood and Moehle (2006), the idealised
“backbone approximates the gradual yielding of the
column with an elastic–perfectly plastic response. Yielding
is assumed to occur once the shear demand reaches the
plastic shear capacity, Vp . For the assumed boundary
conditions... the column plastic shear capacity can be
determined as follows”:
Eq. (1)
In regard to estimating the curvature ductility capacity
of the plastic hinges assumed to form at the top and
bottom of the case study column, NZSEE (2006, sect.
7.2.4) notes that “Priestley and Kowalsky (2000) have
shown that the first yield curvature is given with very good
accuracy as follows… for circular columns”:
Eq. (2)
Other equations are recommended by Priestley and
Kowalsky (2000) for rectangular columns and beams.
Note that first yield curvature can be more accurately
determined by considering a cracked, transformed
cross-section at initial yield of the longitudinal tension
reinforcement. Assuming the column is in double
curvature, fixed against rotation at both ends, and
experiences a linear variation in curvature over its height,
the drift ratio at yield due to flexure can be estimated as
follows:
Eq. (3)
The Elwood and Moehle (2006) backbone model was
developed based on experimental test data comprised
of RC columns with deformed reinforcement bars.
However, all reinforcement in the Art Deco columns
considered by Walsh et al. (2014) was identified as being
comprised of smooth, round bars. Various other studies
into RC columns with similar detailing have found that
round longitudinal bars with ineffective anchorage or
development length can induce failure at relatively low
loads (Fabbrocino et al. 2005), but that properly spliced
or anchored round longitudinal bars are likely to effectuate
higher column deformability than deformed bars (Ricci
et al. 2013) and induce rocking mechanisms at the
base of the columns (Arani et al. 2013). The practicing
engineer should consider closely the anchorage and
lap splice conditions in the column being assessed and
compare the anticipated tensile forces in the longitudinal
reinforcing steel corresponding to plastic flexural capacity
hinge formation to the experimental results reported by
Fabbrocino et al. (2005). The pushover capacity of an
RC column in which longitudinal reinforcement pull-out is
expected to occur prior to plastic hinge formation may be
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81
estimated by considering only the contribution to lateral
resistance from rocking mechanisms as proposed by
Arani et al. (2014).
Where able to be identified on plans, longitudinal
reinforcement bars in the Art Deco columns considered
by Walsh et al. (2014) were either continuous through the
beam-column joints or anchored using 180-degree hooks
with similar geometry and detailing to experimentally
tested RC columns that did not experience pull-out as
reported by Ricci et al. (2013). Hence, for purposes of
estimating Art Deco column capacities, Walsh et al.
(2014) assumed that lap splice and anchorage pull-out
did not govern the drift capacity of any of the columns
considered. For the reported case study column and
in accordance with the recommendation of Arani et al.
(2014) for RC columns with smooth, round longitudinal
reinforcement bars, a bond stress, u , equal to 0.3¥ f'co
(MPa units) was assumed, equalling 1.65 MPa.
According to Elwood and Moehle (2006), “the stress
in the tension reinforcement at the point of effective
yield can be taken equal to the yield stress for columns
with axial load below [N/(Ag'f'co)] = 0.2 and equal to
zero for axial loads above [N/(Ag'f'co)] = 0.5, with a
linear interpolation between these points (Elwood and
Eberhard 2006).” For the case study column, [N/(Ag'f'co)]
= 0.20, and the corresponding stress in the longitudinal
tension reinforcement at the point of effective yield was
determined as fs = 270 MPa.
According to Elwood and Moehle (2006), “Elwood and
Eberhard (2006) have shown that the drift ratio at yield
due to bar slip depends on the column axial load and can
be estimated as follows”:
Eq. (4)
According to Elwood and Moehle (2006), “assuming the
column is fixed against rotation at both ends, the drift
ratio at yield due to shear deformations can be estimated
by idealizing the column as consisting of a homogeneous
material with a shear modulus G.” In accordance with
Elwood and Moehle (2006), the drift ratio at yield due to
shear deformations was determined as follows:
Eq. (5)
According to Elwood and Moehle (2006), “the effective
yield drift ratio can be considered as the sum of the drifts
due to flexure, bar slip, and shear” at yield, resulting in įy
= 0.0089 for the case study column.
According to Elwood and Moehle (2006), “the shear
stress, v , can be estimated based on the plastic shear
capacity” divided by the effective shear area of the
cross-section. In the case of a circular cross-section,
the effective shear area was assumed to be equal to
0.7Ag (Merta and Kolbitsch 2006). Hence, the stress was
determined as follows:
Eq. (6)
Concrete core dimensions were assumed to be
represented by the concrete area bounded by the
outside edges of the transverse reinforcement, consistent
with the assumptions of Mander et al. (1988) and NZS
(2006, sect. 10.1). Hence, for the case study column,
the confined core diameter was determined to be dcore
= 297 mm. The cross-sectional area of the transverse
reinforcing bar was As_trans = 32 mm2 . The volumetric
transverse reinforcement ratio considering the circular
cross-section and spiral reinforcement was determined
to be ȡs = 0.0048 (per Mander et al. 1988). However, the
empirical equations proposed by Elwood and Moehle
(2006) for predicting the column drift ratio at shear failure,
įs , and the column drift ratio at axial failure, įa , were
developed for rectangular columns. Hence, for purposes
of estimating the equivalent rectangular geometry, the
column was transformed into an equivalent rectangular
column using the empirical method proposed by Whitney
(1942). The equivalent column cross-sectional dimension
orthogonal to the considered lateral load was assumed
to be b = 438.8 mm, and the equivalent column cross-
sectional dimension parallel to the considered lateral
load was assumed to be h = 357.6 mm. All longitudinal
reinforcement was assumed equally distributed in one
of two layers relative to the direction of transverse load
considered, with the two layers separated by a distance
equal to 2/3 of the core dimension in the actual octagonal
(circular) cross-section. In the case of the worked
example, this value would be approximately dcore,rect = 200
mm. The width of the confined core in the equivalent
rectangular section was assumed to be equal to the
equivalent width minus the clear cover on each side,
resulting in a value of approximately bcore,rect = 290 mm For
purposes of the worked example, the spiral transverse
reinforcement was treated as equivalent discrete
rectangular hoops vertically spaced centre-to-centre at s
= 90 mm. The resulting transverse reinforcement parallel
to the direction of lateral loading considered for the
equivalent rectangular section was determined to be ȡs,rect
= 0.0025 (per Mander et al. 1988).
In accordance with Elwood and Moehle (2005a, 2006),
the drift ratio at shear failure was determined as follows:
Eq. (7)
Journal of the Structural Engineering Society of New Zealand Inc
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82
In accordance with Elwood and Moehle (2005b, 2006),
the drift ratio at axial failure was determined as follows:
Eq. (8)
Shear strength reduction due to plastic hinging
The Elwood and Moehle (2006) backbone model is
premised on shear failure occurring prior to axial failure.
According to Elwood and Moehle (2006), “experimental
studies have shown that axial load failure tends to occur
when the shear strength degrades to approximately zero
(Yoshimura and Yamanaka 2000). Hence, the final point
on the idealised backbone assumes a shear strength of
zero.” Therefore, in the case that the predicted column
drift ratio at shear failure, įs , is larger than the predicted
column drift ratio at axial failure, įa , (as is the case in
the worked example) then the column’s potential drift
capacity is expected to be governed by the predicted
column drift ratio at shear failure, įs (i.e., the larger value
controls). However, according to Elwood and Moehle
(2006), “before the idealised backbone model can be
used to assess the capacity of an existing reinforced
concrete column, it must be determined whether the
column is expected to experience shear failure after
yielding of the longitudinal reinforcement, a common
characteristic of all the column tests used to develop
the drift capacity models described above. To determine
whether a column is likely to experience shear failure after
flexural yielding, the plastic shear capacity, Vp , should
be compared with an appropriate shear strength model,
such as that proposed by Sezen and Moehle (2004).”
Hence, if shear failure were to occur in a column prior to
flexural yielding, then the column drift ratio at shear failure,
įs , predicted by application of the Elwood and Moehle
(2006) model would likely be overestimated.
The shear capacity provided by the transverse
reinforcement was determined using the equation
recommended by Priestley et al. (1994) for a circular
cross-section as follows:
Eq. (9)
The “shear span” of the column is the distance from the
section of the column sustaining the maximum applied
bending moment under lateral loading to the point of
contraflexure (i.e., section of the column with no applied
bending moment). In a column deformed in double
curvature with assumed flexurally rigid end restraints, the
shear span distance, z , is effectively half of the column
clear-height (e.g., 1965 mm in the case of the worked
example).
As with Elwood and Moehle (2006), the empirical
equation proposed by Sezen and Moehle (2004) for
predicting the column shear strength provided by the
concrete was developed for rectangular columns. The
distance from the extreme compression fibre to the
centroid of the longitudinal tension reinforcement, d , for
the equivalent rectangular section based on the empirical
Whitney (1942) model was determined to be 262 mm.
In accordance with Sezen and Moehle (2004), the shear
capacity provided by the concrete in the equivalent
rectangular section was determined as follows:
Eq. (10)
In accordance with Sezen and Moehle (2004), the
nominal shear strength of the column corresponding
to a displacement ductility demand of 2.0 or less was
determined as follows:
Eq. (11)
To account for shear strength degradation at higher
displacement ductility demands, a reduction factor, k ,
needs to be determined. According to Sezen and Moehle
(2004), “the factor k is defined to be equal to 1.0 for
displacement ductility less than 2, to be equal to 0.7 for
displacement ductility exceeding 6, and to vary linearly for
intermediate displacement ductilities.” In the case of the
worked example, the maximum potential displacement
ductility demand effectuated by plastic hinge development
was determined to be μdispl,col = 3.1 by applying the
predictive method of Elwood and Moehle (2006). Hence,
the strength reduction factor for use in the shear capacity
predictive model of Sezen and Moehle (2004) was
k = 0.92, and the predicted shear strength of the column
corresponding to the ultimate displacement ductility
demand was determined to be kVn = 147 kN.
In interpreting the results of this worked example and
exercises based off of it, Elwood and Moehle (2006)
advise that “...the plastic shear capacity should fall in the
range between the initial shear strength, Vn , and the final
shear strength, 0.7Vn . Because there is some dispersion
between actual and calculated shear strength, some
columns with shear demand less than the calculated
shear, or plastic shear capacity greater than Vn , may still
experience shear failure after flexural yielding. Currently,
engineering judgment is required to select which columns
with Vp outside the Vn to 0.7Vn range are still expected
to experience shear failure after flexural yielding, and
hence can be evaluated using the proposed idealised
backbone model.” In the worked example in which the
minimum axial load was considered, the column shear
ratio at a displacement ductility demand of 2.0 or less,
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83
Vp/ Vn , was determined to be 1.09 and the column shear
ratio at the ultimate displacement ductility demand, Vp
/ kVn , was determined to be 1.18. (These ratios were
altered negligibly when the maximum axial load was
considered and so are not reported here.) Hence, the
case study column would be expected to experience
shear failure just prior to flexural yielding (i.e., be governed
by shear failure criteria per ASCE 2014) and would
be unlikely to actually reach the considered ultimate
displacement ductility demand level of 3.1 corresponding
to flexure-shear failure criteria. For the case study
column, the column drift capacity was assumed equal to
approximately 0.8% where the flexural pushover curve
and shear strength envelope first intersect (see Figure 6).
Figure 6. Shear strength capacity of the case study column
as affected by flexure and shear interaction
(minimum axial load scenario)
Another potentially limiting factor to column drift capacity
is the effect of bidirectional loading on columns with
limited ductility (Boys et al. 2008). Practicing engineers
should use judgment in interpreting the results of
assessment methods based on unidirectional behaviour,
especially when performing a nonlinear time-history
analysis using ground motion records and considering
building spectral responses with relatively strong motions
in two orthogonal directions. Ang et al. (1989) and Wong
et al. (1993) developed a relationship between flexure and
shear interaction based on experimental results of circular
columns to account for the effects of bidirectional loading,
and it was generalised so as to be related to alternative
models for unidirectional loading by Priestley et al. (1994)
as shown in Figure 6.
Flexural strength reduction due to longitudinal bar
buckling
For an alternative column expected to undergo flexural
yielding prior to shear failure, the effect of longitudinal bar
buckling on the flexural strength of the column should
also be considered. An empirical model that can be used
to predict column drift at the onset of longitudinal bar
buckling was proposed by Berry and Eberhard (2005).
This model requires the consideration of two factors
not yet defined in the worked example. The first factor
determined was the effective confinement ratio, which
was defined as follows:
Eq. (12)
The second factor is the transverse reinforcement
coefficient, ke,bb , for which Berry and Eberhard (2005)
recommend that “ke,bb = 40 for rectangular-reinforced
columns and 150 for spiral-reinforced columns…
Because little data were available for large values of
[s /dlong] , ke,bb should be taken as 0.0 for columns in which
[s /dlong] exceeds 6.” The s /dlong ratio for the case study
column was computed to be 3.1. Hence, the transverse
reinforcement coefficient was assumed to be ke,bb =
150. In accordance with Berry and Eberhard (2005), the
estimated drift corresponding to the onset of longitudinal
bar bucking was determined as follows:
Eq. (13)
Note that, for the case study column, the predicted drift
at longitudinal bar buckling far exceeded the predicted
drift at any other limit state (see Figure 6). Hence, bar
buckling was not expected to control the lateral behaviour
of the case study column from the IMS Hastings building.
Column %NBS by displacement-based assessment
The estimated %NBS for the case study column was
determined by dividing the estimated column drift
capacity (assumed as 0.8%, see Figure 6) by the
maximum ULS column drift demand for column E on the
ground storey (0.52%, see Figure 5 and Table 4), resulting
in a %NBS higher than 100. Note that the drift capacity
in the worked example was determined for the interior
ground storey column E whereas the maximum ULS drift
demand for all columns was associated with the exterior
first storey column A (see Figure 5 and Table 4). Hence,
the pushover backbone model analysis would need
to be repeated for different column locations on other
storeys. As noted previously, exterior columns are likely
to experience relatively larger bandwidths of axial loads
effectuated by frame action during earthquake loading
than are interior columns. Two scenarios of flexure and
shear interaction pertaining to minimum and maximum
axial load scenarios, respectively, should be considered
for such columns (see Figure 6).
The ULS column drift demands determined from the
computer-aided NLTHA (see Table 4) were based on
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84
ground-motion time-history records scaled to the ULS
DBE spectrum (i.e., 100%NBS demands, see Figure
3). In order to precisely determine the %NBS for the
columns, however, the time-history records applied in the
NLTHA model should actually be scaled to an increased
target demand spectrum (for potential %NBS scores
above 100%NBS) or a reduced target demand spectrum
(for potential %NBS scores below 100%NBS) until the
maximum drift demands determined from the NLTHA are
equivalent to the estimated drift capacities, at which point
the %NBS would be equivalent to the proportion of the
scaled target spectrum to the ULS DBE spectrum (see
Figure 3).
In the case study assessment presented herein, a
“column sway” collapse mechanism (NZSEE 2006)
was assumed to occur, which generally results in
conservatively high predicted nonlinear column
displacement demands. Also in the case study
assessment presented herein, plastic hinges in the RC
columns were assigned to follow generic FEMA (2000)
and ASCE (2014) criteria, which are included in newer
versions of SAP2000TM by default. Alternatively, the plastic
hinges could also be assigned to behave exactly as
determined by the backbone pushover model proposed
by Elwood and Moehle (2006) while accounting for the
effects of assumed changing axial loads on the columns
at each storey. However, if the nonlinear behaviours
of axial and shear capacities are not considered in
the computational model, then manual calculations to
determine whether shear and axial failure mechanisms
control the collapse of the considered columns are still
required following the computer-aided NLTHA.
Finally, in regard to capacity reduction factors for
assessing RC elements, NZSEE (2006) notes that “the
strength reduction factor ø for flexure should be taken
as 1.0. A strength reduction factor ø for shear of 0.85
should be built into the shear strength.” Hence, at the
discretion of the engineer depending on how conservative
their assumptions for material strengths were, and with
the expectation of shear-controlled failure mechanisms
in some considered RC columns, a reduction in shear
strength of 15% may be warranted.
Effects of URM infill walls on column behaviour
Of the 150 contemporary non-residential buildings
identified in Hastings, at least 40% were identified as
having clay brick URM infill panels (Walsh et al. 2014),
including the IMS Hastings building [see Figure 1(a)–(b)].
In many Hawke’s Bay Art Deco buildings, URM infill
panels only rise partial-height within any given perimeter
frame, truncated most usually by window frames [see
Figure 1(a)]. As noted previously, if the direction of lateral
loading being considered on a given column is in-plane
with partial-height masonry infill, then a “short column”
vulnerability may exist (NZSEE 2014), and the column
failure mechanism is more likely to be shear-controlled
in this direction of loading as the effective shear span
of the column is reduced. Asymmetrically placed infill
panels, typical of buildings which are located on street
corners such as the IMS Hastings building, are capable
of causing plan irregularities and torsional responses, as
simulated in the NLTHA model reported herein. However,
Dowrick and Rhoades (2000) observed that low-rise RC-
framed buildings with asymmetrically placed wall panels
rarely suffered more damage in numerous historic New
Zealand earthquakes than did their counterparts with
symmetrically placed infill walls. Furthermore, Dowrick and
Rhoades (2000) observed that the presence of infill walls
likely benefitted several buildings during earthquakes.
SUMMARY AND RECOMMENDATIONS
The detailed seismic assessment of the Art Deco case
study IMS Hastings building was completed in sequence
with the following generic steps:
• A computer-aided model with assigned plastic hinges
at the tops and bottoms of all columns was subjected
to three time-history records in order to determine
maximum column displacement demands as well as
minimum and maximum column axial loads;
• An empirical nonlinear column pushover model
(Elwood and Moehle 2006) was then utilised in order
to estimate the column displacement capacity under
conditions in which the column would be expected to
flexurally yield prior to failing in shear;
• For the considered column with a high ratio of shear
demand at plastic hinging to nominal shear capacity,
Vp / Vn , additional column failure criteria were
checked against an empirical flexure-shear interaction
model (Sezen and Moehle 2004); and
• A capacity/demand (%NBS) ratio was able to be
determined.
The NLTHA results for the considered case study
indicated that the building would be expected to deform
torsionally in most time-history cases due largely to
eccentrically placed URM infill walls and RC lift shafts.
Nonetheless, the structural redundancy of the building
and contribution from stiffening components considered
in the NLTHA model, including the RC slab and URM infill
walls, would be expected to limit the building’s inter-
storey drifts such that the %NBS of the columns would
remain relatively high. Recommendations exhibited in this
case study for structural engineers performing detailed
seismic assessments on similar interwar Art Deco RC
frame buildings are as follows:
Volume 29 No.1 April 2016
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85
• Owners of regularly configured interwar Art Deco
buildings in Hawke’s Bay may receive higher seismic
assessment scores from engineering consultants
by commissioning invasive and non-invasive
investigations in order to accurately determine material
strengths and structural configurations;
• Engineers seeking accuracy and the avoidance of
excessive conservativeness should focus in their
assessments on leveraging the inherent stiffness
and redundancy of the complete structure to limit
the estimated inter-storey drift demands instead of,
as is traditionally done, focusing on increasing the
estimated strength capacity of structural elements.
Practically, this approach requires that engineers
utilise system-oriented assessment techniques such
as modal response spectrum analyses or, preferably,
displacement-based assessment techniques such as
the nonlinear procedures exemplified herein;
• Potential vulnerabilities identified among the Art Deco
columns that should be closely considered during
assessment include splice or anchorage failure of
smooth reinforcement bars, premature buckling of
longitudinal reinforcement in columns, and column
shear strength degradation at high ductility demands;
and
• Infill walls may have contributed greatly to the
successful performance of similar buildings in
previous earthquakes and should be included as
components in NLTHA models. However, the case
study as presented herein directs the engineer on
how to account for the potentially detrimental effects
caused by infill walls, including the torsional modal
reactions caused by eccentrically placed infill walls
(by identifying the increased displacement demands
on columns located far from the centre of eccentricity
in the NLTHA model) and “short column” behaviour
caused by partial-height infill (by reducing the column
clear-height for pushover capacity modelling).
ADDITIONAL CONSIDERATIONS
The scope of the case study was limited to assessing the
behaviour of a single RC column under lateral loading.
Note that the most significant hazard to people during
an earthquake may not be the failures of load-bearing
structural elements such as RC columns, but rather
the collapse of non-structural parts and components.
Although Hawke’s Bay’s Art Deco building stock has few
tall chimneys and gable end walls, slender RC parapets
and URM infill walls are prominent in Art Deco buildings
(Walsh et al. 2014). These components are prone to out-
of-plane collapse and thus can be especially dangerous
to pedestrians just outside a building (Ingham and Griffith
2011; Cooper et al. 2012; Walsh et al. 2015). In addition,
other building components and behaviours were not
considered in the worked example, including but not
limited to foundation settlement, foundation overturning,
liquefaction, beam shear, slab shear, wall shear, wall
overturning, and beam elongation / diaphragm dilation.
ACKNOWLEDGEMENTS
The authors would like to thank those associated with the
Napier Art Deco Trust (http://www.artdeconapier.com)
for providing the authors with access to the case study
buildings and to building plans considered and referenced
by Walsh et al. (2014). As it relates to the specific case
study reported herein, the authors are especially grateful
to the owners of the IMS Hastings building and to the
structural engineering consultants at EQ STRUC, Ltd.
(http://www.eqstruc.co.nz) who undertook the building
investigation.
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A structural engineer, a chemical engineer, an electrical
engineer and a computer engineer were driving along in a
car when, without warning, the car suddenly stopped.
The structural engineer thought that there was a problem
with the crankshaft.
The chemical engineer thought that there might be a
problem with the petrol.
The electrical engineer thought that there was a problem
with the ignition system.
The computer engineer said, “If we all just get
out of the car and get back in again,
everything should be OK!”
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SESOC Journal
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APPENDIX A: ILLUSTRATED FINDINGS FROM THE INVASIVE INSPECTION OF THE IMS HASTINGS BUILDING
Volume 29 No.1 April 2016
SESOC Journal
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APPENDIX A: ILLUSTRATED FINDINGS FROM THE INVASIVE INSPECTION OF THE IMS HASTINGS BUILDING
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SESOC Journal
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APPENDIX A: ILLUSTRATED FINDINGS FROM THE INVASIVE INSPECTION OF THE IMS HASTINGS BUILDING
